This feature is adapted from two major sources:
It is presented in 6 sections as follows:
You can also find some similar and addition information at the following links in AISC's Modern Steel Construction magazine:
The famous bank robber Willie Sutton was once asked a simple question: why do you rob banks, Willie? His simple response: "because that's where the money is." Sarcastic? Maybe, but his answer showed why he was so successful. The simple question in your mind right now is probably "what does Willie Sutton have to do with steel economy?" Well, like he said,
If you want to save money in steel construction, go where the money is!
To find where the money is, let's take a look at how costs are determined.
When a steel fabricator prepares a cost estimate for a typical project, the following steps are common:
All of the components of the total cost identified in the foregoing estimating process can be classified into one of four categories:
Material costs: This category includes the structural shapes, plates, steel joists,
steel deck, bolting products, welding products, painting products, and any other products
that must be purchased and incorporated into the work. It also includes the waste
materials, such as short lengths of beams (called "drops") that result when
beams are cut to the specified length. By an order of magnitude, the most influential
component of these products on the total material cost of a building structure is the
weight of the structural shapes.
As illustrated in Figure 1, the typical material cost has dropped in recent years from 40 percent of the total cost in 1983 to 26 percent in 1998. This represents a 35 percent decline in material cost over the last 15 years.
Fabrication labor costs: This category includes the fabrication labor required to
prepare and assemble the shop assemblies of structural shapes, plates, bolts, welds and
other materials and products for shipment and subsequent erection in the field. It also
includes the labor associated with shop painting. The total fabrication labor cost is
simply the cost of the shop time required to prepare and assemble these components,
including overhead and profit.
As illustrated in Figure 1, the typical fabrication labor cost has increased slightly in recent years from 30 percent of the total cost in 1983 to 33 percent in 1998. This represents a 10 percent increase in fabrication labor costs over the last 15 years.
Erection labor costs: This category includes the erection labor required to unload,
lift, place and connect the components of the structural steel frame. The total erection
labor cost is simply the cost of the field time required to assemble the structure,
including overhead and profit.
As illustrated in Figure 1, the typical erection labor cost has increased in recent years from 19 percent of the total cost in 1983 to 27 percent in 1998. This represents a 42 percent increase in erection labor costs over the last 15 years.
Other costs: This catch-all category includes all cost items not specifically
included in the three foregoing categories, including outside services other than
erection, shop drawings and the additional costs associated with risk, the need for
contingency, and the schedule requirements of the project.
As illustrated in Figure 1, the typical cost in this category has increased slightly in recent years from 11 percent of the total cost in 1983 to 13 percent in 1998. This represents an 18 percent increase in other costs over the last 15 years.
Figure 1. Material, shop labor, erection labor and other costs; 1983 through 1998.
Obviously, very few projects, designers, fabricators and erectors are exactly alike. Given this, the exact distribution of the total cost among these four categories can and will vary based upon the specific characteristics of a given project, including the design and construction team. In some specialized cases, any one of the four cost centers may dominate the total cost. Nonetheless, it can be stated that the current distribution of cost, rounded to the nearest 5-percent increment, among these four centers for a typical structural steel building is approximately as follows:
Thus, in todays market, labor in the form of fabrication and erection operations typically accounts for approximately 60 percent of the total constructed cost. In contrast, material costs only account for approximately 25 percent of the total constructed cost. Clearly then, least weight does not mean least cost. Instead, project economy is maximized when the design is configured to simplify the labor associated with fabrication and erection. Willie Sutton would go after the labor.
Given The Economy Equation above, the following are 40 basic suggestions that you can use in your office practice today to work smarter, not harder -- and to improve the economy of steel building construction.
With the division of responsibilities for design, fabrication and erection that is normal in current U.S. practice, open communication between the engineer, fabricator, erector and other parties in the project is the key to achieving economy. In this way, the expertise of each party in the process can be employed at a time when it is still possible to implement economical ideas. The sharing of ideas and expertise is the key to a successful project.
Take advantage of a pre-bid conference.
When in doubt about a framing detail or construction practice, consult a knowledgeable fabricator and/or erector. Most will gladly make themselves available at any stage of the game for a pre-bid conference, such as to help with preliminary planning or discuss acceptable and economical fabrication and erection practices. A pre-bid conference can also be used to communicate the requirements and intent of the project to avoid misunderstandings that can be costly. Many times, fabricators and erectors can provide valid cost-saving suggestions that, if entertained, can reduce cost without sacrificing quality.
Issue complete contract documents, when possible.
Design drawings and specifications are the means by which the owner, architect and/or engineer communicates the requirements for structural steel framing to the fabricator and erector. Care in preparing these and other contract documents is important, not only to assure responsive bids or estimates, but also to minimize the potential for misrepresentation, errors and omissions in both bidding and the final product. The most clear, complete and accurate design drawings and specifications will generally net the most accurate and competitive bids. Certainly they are the starting point for economical, timely construction in steel. For guidance on what constitutes complete contract documents, consult the AISC Code of Standard Practice, particularly Section 3 therein. When the nature of the project is such that it is not possible to issue complete contract documents at the time of bidding, clearly provide the scope and nature of the work as far as what the framing will be and what kinds of connections are required.
Don't forget to include the basics.
Show a North arrow on each plan. Show a column schedule. Include "General Notes" that cover the requirements for painting, connections, fasteners, etc. in a manner that is consistent, complementary and supplementary to the specification.
Late details can cost a lot.
Even simple detail items like roof- or floor-opening frames can cost a small fortune if delayed, particularly when the delay forces installation after the steel deck is in place.
Show all the structural steel on the structural design drawings.
As indicated in the AISC Code of Standard Practice, structural steel items should be shown and sized on the structural design drawings. The architectural, electrical and mechanical drawings can be used as a supplement to the structural design drawings, such as by direct reference to illustrate the detailed configuration of the steel framing, but the quantities and sizes should be clearly indicated on the structural design drawings.
Make sure the general contractor or construction manager clearly defines
responsibilities for non-structural and miscellaneous steel items.
Structural and non-structural steel items are identified in AISC Code of Standard Practice Section 2. Many items, such as loose lintels, masonry anchors, elevator framing, and precast panel supports, could be provided by more than one subcontractor. Avoid the inclusion of such items in two bids by clearly defining who is to provide them.
Avoid "catch-all" specification language.
Language like "fabricate and erect all steel shown or implied that is necessary to complete the steel framework" probably sounds good to a lawyer, but it really does not add much to quality or economy because it is nebulous and ambiguous. What is implied? Such language probably results only in arguments, contingency dollars or change orders -- and legal fees.
Avoid language that is subject to interpretation.
Vague notations, such as "provide lintels as required", "in a workmanlike manner", "standard" and "to the satisfaction of the engineer" are subject to widely varying interpretations. Instead, when required, specify measurable performance criteria that must be met.
Use standard tolerances.
ASTM A6/A6M defines standard mill practice. The AISC Code of Standard Practice defines fabrication and erection tolerances. The RCSC Specification covers bolting acceptance criteria. AWS D1.1 establishes weld acceptance criteria. These and other documents provide standard tolerances that are acceptable for the majority of cases. Generally, they present the most efficient practices.
In some cases, more restrictive tolerances may be contemplated for compatibility with the systems and materials that are supported by the structural steel frame. Or tolerances may need to be defined for highly specialized systems or when steel and concrete systems are mated. All non-standard practices should be cost justified.
Specify paint only when it is needed.
Corrosion resistance for architectural or structural purposes may be an important criterion in the performance of structural steel. Often, however, the actual conditions of use do not warrant extensive surface preparation or shop painting, and in these cases no special surface preparation or treatment should be called for. For example, steel that is enclosed in building finishes, fireproofed, or to be in contact with concrete generally need not be painted. Furthermore, if a finish coat is not specified, a shop primer coat need not be specified as it is of only minor influence on corrosion in the construction phase. For more information, see AISC Specification Section M3 and it's Commentary.
When painting is necessary, don't ask too much of the shop coat.
The standard shop coat is usually about one mil thick, but can commonly vary up to two mils thick. Recognizing that it is very difficult to apply more than two mils of dry paint thickness in one coat without runs, sags or drips, it should also be recognized that shop-coat thicknesses in excess of two mils will generally require multiple coats and substantially increase shop painting costs. Ultimately, it must be recognized that the shop coat of paint (primer) is temporary and will provide minimal protection during exposure for steel that is to receive a finish coat in the field. Depending upon the duration and nature of the exposure, the shop coat may last from several weeks to many months. Regardless, the durability of the shop coat should be considered if an extended delay in the application of the finish coat is anticipated.
Also, when painting is necessary, select the right surface preparation and paint
system for the job.
The three most commonly used surface preparations are SSPC SP-2 (hand-tool cleaning), SSPC SP-3 (power-tool cleaning) and SSPC SP-6 (commercial blast cleaning). SSPC SP-2 or SP-3 cleaning is usually satisfactory for an ordinary shop prime coat. If conditions call for a high-performance paint system for long term, low maintenance protection, SSPC SP-6 is more frequently required. When assemblies are to be blast cleaned, consider the limitations on size and length, which vary depending upon the available equipment.
Careful consideration should also be given to specifying a paint system that will satisfy the required degree of corrosion protection. A high-quality paint system (or galvanizing) can be cost-effective or even essential for certain applications, as in open parking structures. In these cases, life-cycle costing should be performed. An alternative to high-quality surface treatments, in normal atmospheric environments, may be ASTM A588 (weathering) steel.
Watch out for primer/fireproofing incompatibility.
For steel that is to receive spray-applied fire-protection, the fire protection manufacturer or applicator should be consulted to determine their recommendations and/or preferences for painted or unpainted steel. If a painted surface is preferred, the paint should be compatible with the fire protection.
Above all, make sure to coordinate with the architect so that the primer and finish
coat are compatible!
The designer should consult with the paint manufacturers to ensure that shop-coat primers and finish-coat paints are compatible. Incompatible coatings are all too common.
When specifying galvanized members, keep the maximum lengths in mind.
Galvanizing dip tanks are generally limited to a member length of 40 ft. Longer members often can be double-dipped, as long as the noticeable zone of overlap between the two dips into the tank is not objectionable.
Clearly state any inspection requirements in the contract documents.
The scope and type of inspection of structural steel should be indicated in the project specification. Make sure that the requirements for inspection are appropriate for the application. For example, the inspection of groove welds that will always be in compression during their service life is probably not required. Also, make sure shop inspection is scheduled so that it does not disrupt the normal fabrication process.
Avoid the use of brand names when specifying common products.
When many manufacturers make a product, or there are acceptably equivalent products, avoid specifying the product by brand name. When it is necessary to indicate a brand name for the purposes of description, be sure it is a current, readily available product. Whenever possible, allow the substitution of an "equal". One excellent example: paint.
Try to avoid them entirely, but when you can't, clearly identify changes and
Changes and revisions that are issued after the date of the contract generally have some cost associated with them. For example, material may have already been ordered, shop drawings may have already been drawn and shipping pieces may have already been fabricated. Thus, it is best to avoid a default reliance on the change and revision process as a means to expedite schedules. However, when changes or revisions are necessary or desirable, they should be clearly identified so that all parties can recognize them and account for them.
Provide meaningful and responsive answers to requests for information.
When the fabricator asks for a design clarification through an RFI, the most prompt and complete response, within the limitations of the available information, will be beneficial to all parties. If the RFI involves information on a shop drawing approval submission, it is best to provide the most specific answer possible. Try to avoid responses such as "architect to supply", "general contractor to supply", or "verify in field".
Use 50 ksi steel in wide-flange member design.
U.S. wide-flange steel shape production today is normally 50 ksi by default. Specifying ASTM A992, ASTM A572 grade 50 or ASTM A529 grade 50 for W-shapes is actually now less expensive than specifying ASTM A36 material, as explained here. Thus, even if deflection, drift, vibration or minimum-size criteria control, the higher-strength material will still often help at little or no added cost for bolt bearing and block shear rupture strength as well as in the elimination of detail material like transverse stiffeners and web doubler plates. For further information, see Carter.
Use 36 ksi steel for plates and angles.
ASTM A36 material is still predominant in angles; so much so that it is difficult to obtain 50 ksi angle material, except by special order from the rolling mill. Additionally, there is still a cost differential between 36 ksi and 50 ksi plate products.
Consider the use of hollow structural sections (HSS).
Square and rectangular HSS are available in ASTM A500 grades B and C with 46 and 50 ksi yield strengths, respectively. Round HSS are available in ASTM A500 grades B and C with 42 and 46 ksi yield strengths, respectively. Although their material cost is generally higher, HSS generally have less surface area to paint or fireproof (if required), excellent weak-axis flexural and compressive strength, and excellent torsional resistance when compared with wide-flange cross-sections.
Be careful when specifying beam camber.
Dont specify camber below ¾-in.; small camber ordinates are impractical and a little added steel weight may be more economical anyway. Also, do not overspecify camber. Deflection calculations are approximate and the actual end restraint provided by simple shear connections tends to lessen the camber requirement. Consider specifying from two-thirds to three-quarters of the calculated camber requirement for beams spanning from 20 to 40 ft, respectively, to account for connection and system restraint. In any case, watch out when rounding up the calculated camber ordinate, particularly with composite designs. Shear studs are unforgiving in that they can protrude through the top of the slab when too little camber is relieved under the actual load. Alternatively, allow sufficient slab thickness to account for reduced actual deflection. For further information, see Ricker.
Another thing to keep in mind: the minimum length of a beam that is to be cambered is about 25 ft. Why? Because the fabrication jig that is used to camber beams is usually configured with pivot restraints that hold the beam from 18 ft to 20 ft apart. To make sure there is adequate beam extending beyond this point to resist the concentrated force from the cambering operation, a 25 ft beam is generally required.
Favor the use of partially composite action in beam design.
Although shear stud installation costs vary widely by region, on average, one installed shear stud equates to 10 lb of steel. Fully composite designs are not usually the most economical because the average weight savings per stud is less than 10 lb. Sometimes, the average weight savings per stud for 50 to 75 percent composite beams can exceed the point of equivalency. In some cases, non-composite construction can be most economical. For further information, see Lorenz and Ruddy. A caveat: make sure that the beam in a composite design is adequate to carry the weight of the wet concrete.
When composite construction is specified, the size, spacing, quantity and pattern of placement of shear stud connectors should be specified. It should also be compatible with the type and orientation of the steel deck used.
When evaluating the relative economy of composite construction, keep in mind that most shear stud connector installers charge a minimum daily fee. So, unless there are enough shear stud connectors on a job to warrant at least a days work, it may be more economical to specify a heavier non-composite beam.
Shear stud connectors should be field installed, not shop installed. Otherwise, they are a tripping hazard for the erector's personnel on the walking surface of steel beams.
Consider cantilevered construction for roofs and one-story structures.
Cantilevered construction was invented primarily to reduce the weight of steel required to frame a roof. Although today we are less concerned with weight savings than labor savings, cantilevered construction may still be a good option. Why? Because the associated connections of the members are generally simple to fabricate and fast and safe to erect. So cantilevered construction is still very much a potential way to save money.
Use rolled-beam framing in areas that will support mechanical equipment.
It always happens. The structural design is performed based upon a preliminary estimate of the loads from the mechanical systems and units. Later, the mechanical equipment is changed and the loads go up -- way up -- sometimes after construction has begun. Rolled-beam framing offers much greater flexibility than other alternatives, such as steel-joist framing, to accommodate these changing design loads.
Optimize bay sizes.
It is still a good idea to design initially for strength and deflection. Subsequently, geometry and compatibility can be evaluated at connections, with shape selections modified as necessary. Ruddy suggested that using a bay length of 1.25 to 1.5 times the width, a bay area of about 1000 ft2, and filler beams spanning the long direction combine to maintain economical framing. But,
Avoid shallow beam depths that require reinforcement or added detail material at end
Detail material such as reinforcement plates at copes and haunching to accommodate deeper, special connections is typically far more expensive than simply selecting a deeper member that can be connected more cleanly. If the beam is changed from a W16x50 to a W18x50, the simplified connection is attained virtually for free. And,
Dont change member size frequently just because a smaller or lighter shape can be
Try to get the usage of any given member size to a mill order quantity (approximately 20 tons). Of course smaller quantities can be used and are commonly purchased by the fabricator from service centers (at a cost premium), but detailing, inventory control, fabrication and erection are all simplified with repetition and uniformity. Keep in mind that economy is generally synonymous with the fewest number of different pieces. This same idea applies when selecting the chords and web members in fabricated trusses. Above all,
Select members with favorable geometries.
Watch out for connections at changes in floor elevations; a deeper girder may simplify the connection detail. Also, watch out for W10, W8 and W6 columns, which can have narrow flanges and web depth; connecting to either axis is constrained and difficult. It is often most helpful to make rough sketches of members to approximate scale in their relative positions to discover geometric incompatibilities.
Use repetitive plate thicknesses throughout the various detail materials in a
Just like with member sizing, the use of similar plate thickness throughout the job is generally more economical than changing thicknesses just because you can. For example, use one or two plate thicknesses for all the column base plates. This same idea applies for other detail materials, such as transverse stiffeners and web doubler plates.
Design floor framing to minimize the perceptibility of vibrations.
Floor vibration can be an unintended result in service when floors are designed only for strength and deflection limit-states and an absolute-minimum-weight system is chosen. Todays lighter construction, when combined with the lack of damping due to partitionless open office plans and light actual floor loadings (in the era of the nearly paperless office), has exacerbated the potential for floor vibration problems. Fortunately, design criteria to prevent perceptible floor vibrations from occurring are available; see Murray et al.
When designing for snow-drift loading, decrease beam spacing as the framing approaches
the bottom of a parapet wall.
Reduced beam spacing allows the same deck size to be used and the same beam size to be repeated into a parapet against which snow may drift. This is generally more economical than maintaining the same spacing and changing the deck and beam sizes.
Minimize the need for stiffening.
When required at locations of concentrated flange forces, transverse stiffeners and web doubler plates are labor intensive detail materials. For the sake of economy, using 50 ksi steel and/or a member with a thicker flange or web can often eliminate them. In the latter case, consider trading some less expensive member weight for reduced labor requirements. Always remember to reduce the panel-zone web shear force by the magnitude of the story shear. This can often mean the difference between having to use a web doubler plate and not. For further information, see Carter.
Economize web penetrations to minimize or eliminate stiffening.
Web penetrations in beams are often a cost-effective means of minimizing the depth of a floor system that contains mechanical or electrical ductwork. However, if they are numerous and require stiffening, it is probably more economical to eliminate them and pass all ductwork below the beams, if possible. Thus, stiffening at web penetrations should be called for only if required. The use of a heavier beam, a relocated opening, a change in the size of the opening, and the use of current design procedures can often eliminate the need for reinforcement of beam web penetrations. If web penetrations are to be use and stiffening is required, the most efficient and economical detail is the use of longitudinal stiffeners above and below the opening as illustrated in Figure 2. For more information, see Darwin.
Figure 2. Web penetration reinforcement of an I-shaped beam.
Eliminate column splices, if feasible.
On average, the labor involved in making a column splice equates to about 500 lb of steel. Consider the elimination of a column splice if the resulting longer column shaft remains shippable and erectable. If a column is spliced, locating the splice at 4 to 5 ft above the floor will permit the attachment of safety cables directly to the column shaft, where needed. It will also allow the assembly of the column splice without the need for scaffolding or other accessibility equipment. If the column splice design requires welding in order to attain continuity, consider the use of PJP groove welds rather than CJP groove welds for economy.
Configure column base details that are erectable without the need for guying.
Use a four-rod pattern, base-plate thickness, and attachment between column and base that can withstand gravity and wind loads during erection. At the same time, make sure the footing detail is also adequate against overturning due to loads during erection. For further information, see Fisher and West. This reference contains minimum column base details for various column heights and wind exposures are recommended. And, ...
Make your column base details repetitive, too.
The possibility of foundation errors will be reduced when repetitive anchor-rod and base-plate details are used. Keep your anchor-rod spacings uniform throughout the job. Use headed rods or rods that have been threaded with a nut at the bottom if there is any calculated uplift. Otherwise, hooked rods can also be used if desired. Be sure to identify both the length of the shaft and the hook if so.
Allow the use of the right column-base leveling method for the job.
Three methods are commonly used to level column bases: leveling plates, leveling nuts and washers and shim stacks and wedges. Regional practices and preferences vary. However, the following comments can be stated in general. Leveling plates lend themselves well to small- to medium-sized column bases, say up to 24 in. Shim stacks and wedges, if used properly, can be used on a wide variety of base sizes. Proper use means maintaining a small aspect ratio on the shim stack, possibly tack welding the various plies of the shim stacks to prevent relative movement and secure placement of the devices to prevent inadvertent displacement during erection operations and when load is applied. Leveling nuts and washers lend themselves well to medium-sized base plates, say 24 in. to 36 in., but are only practical when the four-rod pattern of anchor rods is spaced to develop satisfactory moment resistance. Large column base plates, say over 36 in., can become so heavy that they must be shipped independently of the columns and preset, in which case grout holes and special leveling devices are usually required.
Don't over-specify the details of secondary members.
For example, spandrel kickers and diagonal braces can often be provided as square or bevel-cut elements that get welded into the braced member and structural element that provides the bracing resistance with a very simple line of fillet weld. In contrast, it is very costly to require that such secondary details be miter-cut to fit the profile of a member or element to which it is connected and welded all-around.
Keep relieving angles in a practical size range.
The thickness of relieving angles is normally 5/16 in. or 3/8 in. If a greater thickness is required for strength, the basic design assumptions should be reviewed and perhaps modified. If vertical and/or horizontal adjustment of masonry relieving angles is required, the amount of adjustment desired should be specified and the fabricator should be allowed to select the method to achieve this adjustment, such as by slotting or shimming. Final adjustments to locate relieving angles should be made by the mason, preferably after dead load deflection of the spandrel member occurs.
Consider if heavy hot-rolled shapes are really necessary in lighter and
Ordinary roof openings can usually be framed with angles rather than W-shapes or channels. As another example, heavy rolled angles for the concrete floor slab stop (screed angles) are unnecessary if a lighter gage-metal angle will suffice (something in the 10 to 18 gage range, depending upon slab thickness and overhang). These lighter angles can often be supplied with the steel deck and installed with puddle welding, simplifying the fabrication of the structural steel. Small roof openings on the order of 12 in. square or less probably need not be framed at all unless there is a heavy suspended load, such as a leader pipe.
40 more basic suggestions that you can use in your office practice today to work smarter, not harder -- and to improve the economy of steel building construction.
Limit the use of different bolt grades and diameters.
It is seldom feasible to use more than one or two combinations of bolt diameters and grades on a project. The use of different diameters for different grades simplifies the quality assurance task of ensuring that each strength grade was used in the proper location. It also allows more shop efficiency in the drilling or punching operations.
Use ASTM A325
bolts whenever possible.
They are strong, ductile, and reliable -- the best fastener value.
Limit bolt diameter to 1 in.
Diameters of 3/4-in. and 7/8-in. are preferred and 1-in. diameter is still within the installation capability of most equipment. Larger diameters require special equipment and increased spacings and edge distances than are typical.
Select the right bolt hole type for the application.
In steel-to-steel structural connections, standard holes can be used in many bolted joints and are preferred in some cases. For example, standard holes are commonly used in girder and spandrel connections to columns to more accurately control the dimension between column centers and facilitate the plumbing process. In large joints, particularly those in the field, the use of oversized holes or slotted holes can reduce fit-up and assembly time and the associated costs. For further information, see the RCSC Specification.
Open holes need not be filled for structural purposes.
Sometimes, bolt holes remain in the structure without bolts in them. Whether this is because a temporary bolted connection was made at that location, a design modification was made during construction or for other reasons, there is no structural consequence of not having a bolt installed in the hole. Said another way, the hole has the same effect on the member it pierces whether there is a bolt in it or not.
As another example, consider the connection between a roof purlin that runs over the top of the rafter. It may be possible to punch two diagonal holes in the rafter and four in the purlin, which allows greater repetition of member configuration throughout the job and reduces the chance that the holes may get punched opposite to what is desired, particularly if "opposite hand" members are involved.
Don't confuse the requirements for bolts and bolt holes in steel-to-steel
structural connection with those for anchor rods and anchor-rod holes.
There are many differences between steel-to-steel structural connections and steel-to-concrete anchorage applications, including the following important ones:
Use snug-tightened installation whenever possible.
Snug-tightened installation requirements (the full effort of an ironworker with an ordinary spud wrench that brings the connected plies, but without any specific level of clamping force between them) recognize that most bolts in shear need not be pretensioned; for further information, see the RCSC Specification. The maximum shear strength per bolt is realized while the related installation and inspection costs are minimized. As given in the AISC Specification, the cases that must be pretensioned include: tall building column splices, connections that brace tall building columns, some connections in crane buildings, bolts subject to direct tension (AISC and RCSC are currently considering a relaxation of this requirement for ASTM A325 bolts in non-fatigue and non-impact applications), connections subject to impact or significant load reversal, and slip-critical connections.
Permit the use of any of the four approved methods for pretensioning high-strength
bolts, when pretensioning is necessary.
RCSC provides four approved installation methods for high-strength bolts that must be pretensioned: the turn-of-nut method, the calibrated wrench method, the twist-off-type tension-control bolt method and the direct-tension-indicator method. Different fabricators and erectors have different preferences, which mostly center on the installation cost that is associated with their use of each of these methods. When properly used in accordance with RCSC requirements, these methods all provide acceptable results. Therefore, it is in the interest of economy to allow the installer the flexibility to choose their preferred method and properly use it. For more information, see the RCSC Specification.
Minimize the use of slip-critical connections.
Most connections with three or more bolts have normal misalignments that would cause some of the bolts to be in bearing initially. Furthermore, the normal methods of erection usually cause bolts to slip into bearing during erection. Additional slip beyond this point is usually negligible. Special faying surface requirements add cost due to required masking or use of a special paint system. Furthermore, additional installation and inspection requirements add cost. Therefore, generally limit usage of slip-critical connections to those cases required by AISC/RCSC; these include: connections with oversized holes or slotted holes not perpendicular to load; end connections of built-up members, and connections that share load between bolts and welds. For more information, see the AISC Specification and the RCSC Specification.
Follow RCSC Specification requirements for the use of washers.
The use of hardened washers is often unnecessary; see the RCSC Specification for when they are necessary. However, some fabricators and erectors elect to use a hardened washer under the turned element to prevent galling of the connected material, which would otherwise increase wear and tear on the installation equipment. Lock washers are not intended for use with ASTM A325 or A490 bolts, nor should they be specified or used.
Are your bolt threads automatically excluded from the shear planes of the
For ASTM A325 and A490 bolts, a 3/8-in.-thick ply adjacent to the nut will exclude the threads in all cases when the bolt diameter is 3/4 in. or 7/8 in. A 1/2-in.-thick ply adjacent to the nut will exclude the threads in all cases when the bolt diameter is 1 in. or 1 1/8 in. Use a washer under the nut and you can reduce these minimum ply thicknesses by 1/8 in. Depending upon the combination of grip, number of washers and bolt length, a lesser ply thickness may also work with the threads-excluded condition. For further information, refer to Carter.
Take the easy way out on prying action in bolted joints.
Prying action, the additional tension that results in a bolted joint due to deformation of the connected parts, generally requires detailed calculations to determine the combination of bolt diameter, gage and fitting thickness that is required to transmit the design forces. Simplify the whole process as follows. As a first step, calculate the fitting thickness that is required to reduce the prying action to an insignificant (i.e., negligible) amount. If this thickness is present or can be provided, there is no further need for prying action calculations. If this thickness is excessive or cannot be provided, then use the more detailed calculation approach that matches the proper arrangement of bolt diameter, gage and fitting thickness to transmit the loads with prying action.
Configure welded joints to minimize the weld metal volume.
Each pound of weld metal has a deposition time (and labor cost) associated with it. Therefore, the most economical welded joint will generally result when weld metal volume is minimized. Furthermore, reducing the weld metal volume reduces the heat input and the resulting shrinkage and distortion. Minimizing the weld metal also minimizes the potential for weld defects.
Favor fillet welds over groove welds.
Fillet welds generally require less weld metal than groove welds. Additionally, the use of fillet welds virtually eliminates base metal preparation, which is labor intensive.
Configure fillet weld length to reduce weld size.
Compare a 1/4-in. fillet weld 12-in. long with a 1/2-in. fillet weld 6-in. long. These welds have equal strength, but the latter weld requires twice as much weld metal volume and cost. With due consideration of the implications of increased weld length on the size of connecting elements, such as gusset plates, the best balance can be found.
Keep fillet weld sizes at or below 5/16-in., when possible.
Per AWS D1.1, this weld size can be deposited in one pass with the shielded metal arc welding (SMAW) process in the horizontal and flat positions. Larger weld sizes will require multiple passes.
Don't always weld on both sides of a piece just because you can.
There are many applications where it may be possible to weld on one side of a joint only. For example, the attachment of a column base plate to a column can in many cases be made with fillet welds on one side of each flange and the web. This same idea is also sometimes possible with transverse stiffeners, bearing stiffeners and other similar elements.
Use intermittent fillet welds when possible.
Under typical loading, intermittent fillet welds can often be specified and the weld metal volume can be reduced accordingly. However, in applications that involve fatigue, intermittent fillet welds are not permitted.
Favor the horizontal and flat positions.
These positions use the base metal and gravity to hold the molten weld pool in place, allowing easier welding with a faster deposition rate and generally higher weld quality.
Recognize the increased strength in transversely loaded fillet welds.
It has long been known that a fillet weld loaded transversely is up to 50 percent stronger than the same weld loaded longitudinally. AISC Specification Appendix J2.4 provides a means to take advantage of this strength increase [Lesik and Kennedy]. As a result, values in the eccentrically loaded weld group tables in the current AISC Manual are typically from ten to 30 percent higher than in the previous edition. This is particularly useful for transverse stiffener end welds, which are purely transversely loaded and qualify for a full 50 percent increase in shear strength.
Avoid the use of the weld-all-around symbol.
This welding is excessive in most cases. It may even be wrong in some. Welding all around may violate the AISC Specification requirement that welds on opposite sides of a common plane be interrupted at the corner (i.e., when the weld would have to wrap around the corner at an overlap). Also, the weld-all-around symbol should not be used if the entire perimeter of the weld cannot be reached.
Consider the AWS acceptance criteria when an undersized weld is discovered.
AWS D1.1 allows a 1/16-in. undersize to remain if it occupies less than ten percent of the weld length. This recognizes that an attempted repair may create a worse condition than the undersized weld.
Avoid seal welding unless it is required.
Seal welding is generally unnecessary, unless the joint is required to be air-tight or water-tight.
For groove welds, favor partial-joint-penetration (PJP) over complete-joint-penetration
PJP groove welds generally require less base metal preparation and weld metal. They also reduce heat input, shrinkage, and distortion. It is sometimes possible to increase plate thickness and use a PJP groove weld instead of a CJP groove weld.
Select a groove-welded joint with a preparation that minimizes weld metal volume.
Depending upon thickness, a particular combination of root opening and bevel angle will minimize weld metal volume. The combination that requires the least amount of weld metal should be selected. Also, consider double-sided preparation. In some cases, the additional labor to prepare the surface can be offset by savings in weld metal volume (and labor).
Watch out for weld details that will likely cause distortion as the welds cool and
Welding (and in many cases, flame cutting) causes distortion because the heated regions are restrained by the rest of the steel as they cool and contract. Under extreme circumstances, a structural member could be distorted so severely that straightening of the member would be required, particularly in an application that involves architecturally exposed structural steel, with the resultant increase in cost. The potential for distortion frequently can be reduced through proper selection of the connection configuration and joint details. The fabricator is the best source of guidance and advice for avoiding potentially troublesome details.
Know what to look for when monitoring interpass temperatures in welded joints.
For a given level of heat input, the interpass temperature in a weldment is largely dependent upon the cross-sectional area of the element(s) being welded. The larger the area, the faster the heat will be drawn from the weldment. As a rule of thumb, if the cross-sectional area of the weld is equal to or greater than 40 in.2, the minimum interpass temperature should be monitored. Conversely, if the cross-sectional area of the weld is equal to or less than 20 in.2, the maximum interpass temperature should be monitored.
Avoid welding on galvanized surfaces.
When at all possible, avoid design situations that will require welding on galvanized surfaces, particularly in the shop. Special ventilation must be provided in the shop to exhaust the toxic fumes that are produced. Additionally, the galvanizing must commonly be removed by grinding in the area to be welded. This requires the subsequent touching up with cold galvanizing compound after welding and cleaning. All of these operations add cost.
Consider the fabricator's and erector's suggestions regarding connections.
To a large extent, the economy of a structural steel frame depends upon the difficulty involved in the fabrication and erection, which is a direct function of the connections. The fabricator and erector are normally in the best position to identify and evaluate all the criteria that must be considered when selecting and detailing the optimum connection, including such non-structural considerations as equipment limitations, personnel capabilities, season of erection, weight, length limitations and width limitations. The fabricator will also know when variations in bolt diameters and holes sizes, broken gages, combination of bolting and welding on the same shipping piece will incur excessive and costly material handling requirements in the shop.
Design connections for actual forces.
Or at least do not overspecify the design criteria. In U.S. practice, the Engineer of Record sometimes specifies standard reactions for use by the connection designer. These standard reactions can sometimes be quite conservative; look at the extreme example illustrated in Figure 3.. However, design for the actual forces generally allows more widespread use of typical connections, which improves economy. Axial forces, shears, moments and other forces should be shown as applicable so that proper connections can be made and costly overdesign, as well as dangerous underdesign, can be avoided. This applies to shear connections, moment connections, bracing connections, column splices -- all connections! The actual reactions are quite important for the proper design of end connections for beams in composite construction.
Figure 3. Extreme example of what arbitrary connection criteria can mean.
Use one-sided shear connections, when possible.
One-sided connections such as single-plates and single-angles have well-defined performance, are economical to fabricate and are safe to erect in virtually all configurations. When combined with reasonable end-reaction requirements, one-sided connections can be used quite extensively to simplify construction. Sometimes, however, end reactions are large enough to preclude their use because of the strength limitations of such connections.
Avoid through-plates on HSS columns; use single-plate shear connections whenever
A single-plate connection can be welded directly to the column face in all cases where punching shear does not control and the HSS is not a slender-element cross-section. See the AISC HSS Connections Manual.
Consider partially restrained (PR) moment connections.
PR moment connections can provide adequate strength and stiffness for many buildings, particularly those with long frame lines where many connections of reduced stiffness can be mobilized. PR connections can be configured to minimize field welding, simplify the connection details, and speed erection. Engineers can obtain guidance on this subject from a number of sources, including Leon et al.
Design columns to eliminate web doubler plates (especially) and transverse stiffeners
(when possible) at moment connections.
The elimination of labor-intensive items such as web doubler plates and stiffeners is a boon to economy. One fillet-welded doubler plate can generally be equated to about 300 lb of steel; one pair of fillet welded stiffeners can generally be equated to about 200 lb of steel. Additionally, their elimination simplifies weak-axis framing. For further information, see Carter.
The Economy Equation, Ways to Save Time and Money and More Ways to Save Time and Money are almost entirely upon currently established design and construction practices. The following recent developments also have great potential to improve the economy of steel building construction.
Snug-tight bolts in tension applications
The economic benefits of snug-tight bolts can currently be realized in the majority of shear/bearing connections in building structures. However, the current RCSC and AISC Specifications require all bolts in tension applications to be pretensioned. Murray and Johnson showed that the pretension is not critical to performance for ASTM A325 bolts in applications involving tension but not fatigue or impact. Accordingly, it is anticipated that both the RCSC and AISC Specifications will permit snug-tight ASTM A325 bolts in such applications. Thus, the potential for economy in bolted connections will be increased due to the relaxation of the installation and inspection requirements.
Use of one-sided shear connections
The recent development of detailed design procedures for one-sided connections, coupled with erection safety considerations, has led to a more widespread usage of one-sided shear connections, such as single-plate, single-angle and tee shear connections.
Extended end-plate moment connections for seismic loading
With all welding done in the shop with better quality control and lower cost and only bolting in the field, the advantages of extended end-plate moment connections have long been known. However, these connections have historically been limited to use in static loading applications and non- and low-seismic applications.
Meng and Murray showed that, if the design procedure is modified slightly to account for the effects of prying action and weld access holes are not used to make the beam-flange-to-end-plate welds, extended end-plate moment connections are suitable for high-seismic loading. The use of weld access holes causes an increase in the strain rate in the beam flange under load and results in a premature flange fracture.
At the same time, improvements in cutting and drilling equipment have made large moment end-plate connections an increasingly viable alternative for moment resisting frames. Current cutting methods can produce an acceptably square beam end without milling. Additionally, computer-controlled drilling equipment can produce accurate hole placement in both the end plates and the column flanges, resulting in excellent alignment for field assembly.
Other Seismic Moment Connections
Tremendous advancements have also been made in the inelastic deformation capabilities of welded beam-to-column moment connections for seismic applications. At the same time, more stringent performance requirements have been implemented in the AISC Seismic Provisions. Several connection alternatives that can withstand inelastic rotations of at least 3 percent have been developed using reinforcement, including cover plates, ribs and haunches [FEMA and FEMA]. Additionally, the reduced beam section (dogbone) approach has also demonstrated excellent inelastic performance [Iwankiw, Engelhardt and Carter and Iwankiw]. At the same time, bolted solutions, such as extended end-plates (see above), flange tees and flange plates are being investigated further [Meng and Murray, FEMA and FEMA]. Some proprietary alternatives also have been developed. Further information on seismic moment connections is also available in Engelhardt and Sabol.
Research is continuing in this area through the SAC Joint Venture, a partnership formed by the Structural Engineers Association of California (SEAOC), Applied Technology Council (ATC) and California Universities for Research in Earthquake Engineering (CUREe).
Connections Involving HSS
The AISC HSS Connections Manual has drawn together information on connections for HSS from a variety of sources. It includes specific coverage of shear, moment, bracing, truss and other connections as they are commonly found in tubular construction for buildings.
Framing for Reduced Floor-to-floor Heights
The primary design characteristic for some buildings is a reduced floor-to-floor height. Although this characteristic has often led to construction in reinforced concrete in the past, steel systems are increasingly common. Some layouts have utilized conventional framing to achieve a floor-to-floor height as small as 8 ft-8 in. with the structure well integrated into the partitions and other architectural features. Other framing systems such as the staggered truss system [Cohen] have also been used to achieve such a reduction.
Several promising recent advancements also deserve mention. In the United Kingdom, a system called Slimflor has been developed, wherein the infill beams are nested inside and supported by the wider bottom flange of an asymmetrical girder [Lawson et al.]. Spans in the range of 20 to 30 ft are practical for very shallow framing depths. Additionally, in the United States, a preliminary study by Murray on a two-way composite floor system spanning 30 to 40 ft between similar asymmetrical girders indicates the potential for a very competitive system. The two-way composite floor system illustrated in Figure 4 is constructed with deep metal deck spanning one direction, a very shallow form deck spanning the other direction, and double-headed self-tapping screws, which interconnect the two plies of metal deck and achieve composite action with the topping concrete. For more information, see Hillman and Murray, Hillman and Murray and Hillman and Murray.
Figure 4. A two-way steel-deck composite floor system.
Composite "joists", which are actually light trusses with composite action, have recently been introduced. This type of construction allows for economical ultra-clear span construction, and has been used in applications with spans from 45 to 120 ft. Ductwork and other mechanical system components can pass through the open-web trusses, allowing for reduced building height without the need to create web penetrations as for solid-web members. In-service measurements of floor accelerations have shown that floor vibrations are not a significant problem in these systems. For more information, see Easterling et al., Easterling et al. and Murray.
The Uniform Force Method for Bracing Connections
Until recently, there has not been a systematic approach to the analysis and design of bracing connections, such as the connection shown in Figure 5. There is now available just such a method which is based on analytical and experimental results and not on the usual simple beam and strut formulas applied to various cut sections. The method is called the Uniform Force Method and has been adopted by AISC for use in the AISC Manual.
Figure 5. A typical bracing connection.
The admissible force distribution for this method is shown in Figures 6 and 7. The force distribution is called admissible in the sense of the lower bound theorem of limit analysis because it satisfies equilibrium for the free body diagrams shown in Figures 6 and 7, i.e., the gusset in Figure 6 and the beam and column and Figure 7, with absolutely no additional forces required anywhere.
Figure 6. Force distributions in gusset plates per the Uniform Force Method.
Figure 7. Force distributions in the beam and column per the Uniform Force Method.
The idea for the coincident system of forces shown on the gusset free body diagram of Figure 6 comes from the work of Richard, who showed that the force resultants on the gusset edges fall within the regions shown cross-hatched in Figure 8. Each cross-hatched region of Figure 8 contains the resultants for six cases in which the connections of the gusset to the beam and column were varied from bolted to welded. It can be seen by comparing Figure 8 with Figure 6 and the uniform force method captures the essence of Richards results.
Figure 8. Gusset edge force resultant envelopes for
45-degree working point models (adapted from Richard).
In order to validate the method, it was used to predict the failure load and controlling limit state of six full-scale test specimens for which the actual failure load controlling limit are known. A typical test specimen of Bjorhovde and Chakrabarti is shown in Figure 9 and that of Gross and Cheok is shown in Figure 10. Table 1 shows the limit states that have been identified for these specimens, and Table 2 shows the limit states applied to each connection interface.
Figure 9. Test configuration used by Bjorhovde and Chakrabarti.
Figure 10. Test configuration used by Gross and Cheok.
Table 3 shows the results of analyzing the six test specimens by the uniform force method, and comparing the results with the actual physical test results. Predicted and actual connection failure interfaces, limit states, and failure loads are compared. In all cases, the uniform force method gave a conservative prediction of the connection capacity. The first three tests, those of Bjorhovde and Chakrabarti, show extremely close correspondence between the predictions and the physical tests. The second three tests, those of Gross and Cheok, show that the predictions of the uniform force method underestimate the test loads, that is, the uniform force method gives a conservative prediction of capacity. There are two reasons for this.
As Gross points out, his tests include gusset buckling, and the current method for gusset buckling, treating the gusset as a strip column, is known to be conservative because it ignores the elastic support provided by the continuity of the gusset plate. Bjorhovdes tests included only a brace tensile load, so this effect was not addressed.
The second reason involves the distortion of the frame which is ignored by the uniform force method. Gross tests included this but Bjorhovdes did not. When the frame distorts, the forces on the connection interfaces undergo a redistribution somewhat analogous to what happens during shakedown in plastic analysis. The resulting internal force distributions are more benign than those predicted based on pure equilibrium considerations and result in increased load carrying capacity. A discussion of this phenomenon is given by Thornton and Kane.
Table 1. Limit-states identification for bracing connections
|Bolt shear fracture||1|
|Bolt shear/tension fracture||2|
|Gross section yielding||7|
|Net section fracture||8|
|Fillet weld fracture||9|
|Beam web yielding (beyond k-distance)||10|
|Bending yielding (including prying action)||11|
|Bending fracture (including prying action)||12|
|Table 2. Limit-states considered for each interface of bracing connections|
|Connection interface (note 1)||Connection element||Limit states (note 2)|
|Brace-to-gusset (A)||Bolts to gusset||1|
|Gusset||3, 4, 5, 6|
|Bolts to brace||1|
|Brace||5, 6, 7, 8|
|Splice plates for WT's||5, 6, 7, 8|
|Gusset-to-column (C)||Bolts to gusset||1|
|Fillet weld to gusset||9|
|Gusset||6, 7, 8|
|Bolts to column||2|
|Clip angles||6, 7, 8, 11, 12|
|Column||6, 11, 12|
|Beam-to-column (D)||Bolts to beam web||1|
|Fillet weld to beam web||9|
|Beam web||6, 7, 8|
|Bolts to column||2|
|Clip angles||6, 7, 8, 11, 12|
|Column||6, 11, 12|
|Table 3. Comparison of Uniform Force method predicted results with test results.|
|Test specimen||Predicted results (note 1)||Test results (note 1)||Test
divided by predicted
|A, kips||B, kips||C, kips||D, kips||Strength, kips||Controlling interface:||Strength, kips||Controlling interface:|
|Bjorhovde/ Chakrabarti 30º||142
|Bjorhovde/ Chakrabarti 45º||142
|Bjorhovde/ Chakrabarti 60º||142
|-- (note 2)||77
|-- (note 2)||84
|(1) Numbers in
parentheses below the strengths in this table are limit-state numbers as given in Table 1.
(2) No limit. This part of the connection does not carry any of the brace load.
In the opinion of the authors, Charles J. Carter, Thomas M. Murray and William A. Thornton, the following trends and needs are critical to the continued advancement of the state-of-the-art in steel design and construction.
Simplification of Analysis Requirements
Although consideration of second-order effects in the structural analysis is a Specification requirement in all cases, the actual impact of second-order effects is negligible in many practical cases. Most building structures in the United States are not taller than four stories and the actual increase in moments due to deformations of the frame is in many practical cases negligible. At the same time, there are also cases where the consideration of second-order effects is quite important. Some work has already been done [ASCE], but additional simplification of analysis requirements, particularly the Specification requirements, is needed.
Composite Framing Systems
It would be advantageous for the steel design community and construction industry to innovate with practical systems that marry structural steel and reinforced concrete for each of their benefits. For example, a system utilizing composite connections between steel floor framing and concrete-encased steel columns (also known as steel-reinforced-concrete) or reinforced concrete columns offers a decided advantage because of all the deflection problems and structural inefficiencies associated with concrete-slab floor-framing. Such hybrid systems would provide increased column spacing flexibility and improved control of floor deflection and vibration characteristics. Additionally, for seismic design applications, these systems would provide increased structural damping and improved structural ductility.
Development of More Advanced Column Stiffening Design Procedures
Current Specification provisions for the design of transverse stiffeners and web doubler plates do not consider that the presence of the web doubler plate may eliminate the need for the transverse stiffeners. The interaction between these elements and effect of the presence of one on the demands on the other should be investigated. Additionally, alternative reinforcement approaches to costly web doubler plates should also be investigated.
Development of a Design Approach for Fire
The current and conventional approach to treatment of fire in a steel structure is to follow prescriptive guidelines for the protection of the structural system against the heat generated by the fire. The actual demands on the structural system in a fire are known to be quite conservative in relation to what the prescriptive system provides protection against. Because fire protection is a significant cost associated with steel construction, the progression toward a less prescriptive system based more on design and engineering is unavoidable. Recent testing in the United Kingdom and elsewhere as reported in Dowling and Burgan emphasizes that a shift toward engineered fire protection would have a significant and beneficial impact on the economy of structural steel buildings.
Note: many of the following references are listed in our feature Technical Bibliography.
American Institute of Steel Construction, Code of Standard Practice for Steel Buildings and Bridges, June 10, 1992, AISC, Chicago, IL, USA, 1992.
American Institute of Steel Construction, Hollow Structural Sections Connections Manual, AISC, Chicago, IL, USA, 1997.
American Institute of Steel Construction, Load and Resistance Factor Design Specification for Structural Steel Buildings, December 1, 1993, AISC, Chicago, IL, USA, 1993.
American Institute of Steel Construction, Load and Resistance Factor Design Manual of Steel Construction, AISC, Chicago, IL, USA, 1994.
American Institute of Steel Construction, Seismic Provisions for Structural Steel Buildings, AISC, Chicago, IL, USA, 1997.
American Society for Testing and Materials, ASTM A6/A6M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A36/A36M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A325/A325M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A354/A354M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A449/A449M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A490/A490M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A500, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A529/A529M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A588/A588M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A572/A572M, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM A992, ASTM, Conshohocken, PA, 1997.
American Society for Testing and Materials, ASTM F1554, ASTM, Conshohocken, PA, 1997.
American Society of Civil Engineers, Effective Length and Notional Load Approaches for Assessing Frame Stability: Implications for American Steel Design, ASCE, Reston, VA, USA, 1997.
American Welding Society, AWS D1.1 Structural Welding CodeSteel, AWS, Miami, FL, USA, 1998.
Bjorhovde, Reidar and Chakrabarti, S.K., "Tests of Full Size Gusset Plate Connections," ASCE Journal of Structural Engineering, Vol. 111, No.3, March, pp. 667-684, ASCE, Reston, VA, 1985.
Carter, C.J., AISC Design Guide #13 Wide-Flange Column Stiffening at Moment Connections: Wind and Seismic Applications, AISC, Chicago, IL, 1999.
Carter, C.J., "Are You Properly Specifying Materials?," Modern Steel Construction, AISC, Chicago, IL. Part 1 -- Structural Shapes (January 1999 issue); Part 2 -- Plate Products (February 1999 issue); Part 3 -- Fastening Products (March 1999 issue).
Carter, C.J., "Specifying Bolt Length for High-Strength Bolts," Engineering Journal, 2nd Quarter, AISC, Chicago, IL, USA, 1996.
Carter, C.J. and Iwankiw, N.R., "Improved Ductility in Seismic Steel Moment Frames with Dogbone Connections," Journal of Constructional Steel Research, April-June, Elsevier Science Ltd, Kidlington, Oxford, UK, 1998.
Cohen, M.P., "Design Solutions Utilizing the Staggered-steel Truss System," Engineering Journal, 3rd Quarter, AISC, Chicago, IL, USA, 1986.
Darwin, D., AISC Design Guide 2 Steel and Composite Beams with Web Openings, AISC, Chicago, IL, USA, 1990.
Dowling, P.J. and Burgan, B.A., "Steel Structures in the New Millennium," Journal of Constructional Steel Research, April-June, Elsevier Science Ltd, Kidlington, Oxford, UK, 1998.
Easterling, W. S., D. R. Gibbings, and T. M. Murray, "Strength of Shear Studs in Steel Deck on Composite Beams and Joists", Engineering Journal, American Institute of Steel Construction, Vol. 30, No. 2, 2nd Qtr., pages 44-55, AISC, 1993.
Easterling, W. S., D. R. Gibbings and T. M. Murray, "Composite Beams and Joists", Structural Engineering in Natural Hazards Reduction, Proceedings of papers presented at the ASCE Structures Congress '93, Irvine, CA, April 19-21, 1993, pages 1257-1260, ASCE, 1993.
Engelhardt, M.D., Test Reports on Curved Dogbone, University of TexasAustin, Austin, TX, USA, 1996.
Engelhardt, M.D. and Sabol, T.A., "Seismic-Resistant Steel Moment Connections," Progress in Structural Engineering and Materials, Vol. 1, No. 1, CRC Ltd, Watford, Hertfordshire, UK, 1997.
Federal Emergency Management Agency, FEMA 267 Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures, FEMA, Washington, DC, USA, 1995
Federal Emergency Management Agency, FEMA 267A Interim Guidelines Advisory No. 1, Supplement to FEMA 267, FEMA, Washington, DC, USA, 1997
Fisher, J.M. and West, M.A., AISC Design Guide 10 Erection Bracing of Low-Rise Structural Steel Frames, AISC, Chicago, IL, USA, 1997.
Gross, John L., "Experimental Study of Gusseted Connections," AISC Engineering Journal, 3rd Qtr. Vol. 7, pp. 89 - 97, AISC, Chicago, IL, 1990.
Gross, John and Cheok, Geraldine, "Experimental Study of Gusseted Connections for Laterally Braced Steel Buildings," National Institute of Standards and Technology Report, NISTIR 89 - 3849, Gaithersburg, MD, NIST, November, 1988.
Hillman, J.R. and T.M. Murray, "Innovative Floor Systems", Proceedings, 1990 National Steel Construction Conference, pp. 12-1 to 12-27, AISC, Chicago, IL, 1990.
Hillman, J.R. and T.M. Murray, "Innovative Lightweight Floor Systems for Steel Framed Buildings", Mixed Structures, Including New Materials IABSE Symposium, Brussels 1990, IABSE Report No. 60, Zurich, 1990, pages 671-676, IABSE.
Hillman, J.R. and T.M. Murray, "An Innovative Cold-Formed Floor System," Recent Research and Developments in Cold-Formed Steel Design and Construction, Proceedings of the Twelfth International Specialty Conference on Cold-Formed Structures, St. Louis, October 18-19, 1994, pp. 513-522.
Iwankiw, N.R., "Ultimate Strength Considerations for Seismic Design of the Reduced Beam Section (Internal Plastic Hinge)," Engineering Journal, 1st Quarter, AISC, Chicago, IL, USA, 1997.
Johnson, D., Unpublished study by Johnson, Butler Research, Grand View, MO, USA, 1996.
Lawson, R.M., Mullett, D.L. and Rackham, J.W., "Design of Asymmetric Slimflor Beams," Journal of Constructional Steel Research, April-June, Elsevier Science Ltd, Kidlington, Oxford, UK, 1998.
Leon, R.T., Hoffman, J.J, and Steager, T., AISC Design Guide 9 Partially Restrained Composite Connections, AISC, Chicago, IL, USA, 1996.
Lesik, D.F. and Kennedy, D.J.L., "Ultimate Strength of Fillet Welded Connections Loaded in Plane," Canadian Journal of Civil Engineering, Vol. 17, No. 1, National Research Council of Canada, Ottawa, Canada, 1990.
Lorenz, R.F., "Some Economic Considerations for Composite Floor Beams," Engineering Journal, 2nd Quarter, AISC, Chicago, IL, USA, 1983.
Meng, R.L. and Murray, T.M., "Seismic Performance of Bolted End-plate Moment Connections," Proceedings of the1997 AISC National Steel Construction Conference, AISC, Chicago, IL, USA, 1997.
Murray, T. M., "Vibration of Open Web Joist Supported Floor Systems", Proceedings of the AISC National Steel Construction Conference, New Orleans, Louisiana, April 1-3, 1998, Pages 27-1 through 27-18, AISC, Chicago, IL, 1998.
Murray, T.M., "Use of Snug-Tightened Bolts in End-Plate Connections", Proceedings of the Second International Workshop on Connections in Steel Structures, Pittsburgh, PA, April 10-12, 1991.
Murray, T.M., Allen, D.E. and Ungar, E.E., AISC Design Guide 11 Floor Vibrations Due to Human Activity, AISC, Chicago, IL, USA, 1997.
Research Council on Structural Connections, LRFD Specification for Structural Joints Using ASTM A325 or A490 Bolts, AISC, Chicago, IL, USA, 1994.
Richard, Ralph M., "Analysis of Large Bracing Connection Designs for Heavy Construction," Proceedings of the AISC National Engineering Conference, Nashville, TN, June, pp. 31-1 - 31-24, AISC, Chicago, IL, 1986.
Ricker, D.T., "Cambering Steel Beams," Engineering Journal, 4th Quarter, AISC, Chicago, IL,USA, 1989.
Ruddy, J., "Economics of Low-Rise Steel-Framed Structures," Engineering Journal," 3rd Quarter, AISC, Chicago, IL, USA, 1991.
The Society for Protective Coatings, SSPC SP-2, SSPC
The Society for Protective Coatings, SSPC SP-3, SSPC
The Society for Protective Coatings, SSPC SP-6, SSPC
Thornton, W.A. and Kane, T., Connections, Chapter 7 of Steel Design Handbook, Edited by A.R. Tamboli, McGraw Hill, New York, pp. 7-55 - 7-60, 1997.
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